Good morning! This is Back of the Envelope – the place to learn structural engineering in tiny bites .

In today’s article, I will talk about how to do a quick-n-dirty preliminary check to make sure your composite steel deck meets the fire rating requirement.

This is something that confuses me almost all the time… but there are some rule of thumb that could be applied for back-of-the-envelope checks.

Let’s dive in.

(Estimated read time = 1 minute and 30 seconds — I told you, tiny bites)

By the way, this is a rehash of an article I wrote in my weekly newsletter, “Back of the Envelope” — where I teach you SE-related things in 5 minutes (or less), once a week.

What is the fire rating requirement?

The first step is to obtain the structure’s required fire rating from the client/architect. This is based on the type of building and other goodies per IBC Table 601.

Spray-applied fire resistive material underside of deck?

Once you know the required rating, you should then find out if “spray-applied fire resistive materials” (SFRM) will be applied to the underside of the deck.

(Some people refer to the SFRM using the product name Monokote).

Rule of thumb

Now here comes the rule of thumb:

If SFRM will be applied to the underside of the deck, then the concrete thickness above the deck would generally be 2-1/2” thick.

And it can be either normal weight or lightweight.

You’ll achieve a 1 to 4-hour rating with most UL assemblies (the architect and/or fire protection engineer need to detail that).

(source: Verco Deck Binder)

On the other hand, if SFRM will NOT be applied to the underside of the deck, then the thickness of the concrete varies depending on the required rating.

The legacy Verco catalog had this handy table below (it’s a rough generalization of all the UL assemblies):

(source: legacy Verco catalog)

For example, the thinnest configuration to get a 1-hour rating would be using 2-1/2” lightweight concrete over 1-1/2” deck (watch out for unshored span though – topic for another email).

A very common 2-hour rating configuration that I have seen is 3-1/4” lightweight concrete over 3” deck.

And there you have it. Hope that all makes sense.

Now you should be able to come up with a preliminary concrete thickness based on the required rating.

You could then move on to figure out the other design requirements (e.g., unshored span, vertical capacity, diaphragm capacity…etc.). We'll save those for another email.

Thanks for reading and enjoy the rest of your week!

The post Rule of thumb for concrete thickness (over steel deck) to meet fire rating requirement appeared first on Structural Engineer HQ.

In today’s post, I will quickly summarize what I learned from a recent webinar called “Common Challenges in Wood Lateral System Layout.” It was presented by Terry Malone, Senior Technical Director at WoodWorks, who has decades of experience designing structures.

The webinar’s goal was to help us recognize things that we need to watch out for during schematic design (so that we can be prepared to face the challenges… or propose alternatives to the client).

Let’s dive in:

•  Terry’s thought process
•  Common challenges
•  My non-technical takeaway

(Estimated read time: 4 minutes and 10 seconds )

By the way, this is a rehash of an article I wrote in my weekly newsletter, “Back of the Envelope” — where I teach you SE-related things in 5 minutes (or less), once a week.

If you enjoyed reading it, consider subscribing at the end of the post to be one of the first to get new emails every Thursday (which I eventually rehash onto LinkedIn and Structural Engineer HQ a couple of weeks later).

Terry’s thought process

Terry has a standard process that he goes through when looking at a wood building:

1. First, using PDF markups, overlay the shear walls from floor to floor so you can see how they stack (or not). Use different colors to represent shearwalls on different floors.
2. Next, locate the diaphragm boundaries to get a sense of the load paths.
3. Based on what you see, identify potential irregularity and load path challenges, such as offset walls, large openings, cantilever diaphragms…etc.
4. Lastly, form opinions and develop possible solutions based on what you’ve identified.

Here is an example of what that pdf might end up looking like:

(Source: WoodWorks/Terry – link at the end)

In a nutshell, step back to look at the building from a global perspective before pulling out your calculator to crunch the numbers.

This will help you better understand how the building might behave or perform and help you recognize what you could potentially improve.

Common challenges

Now, here are some of the lateral design challenges that come up frequently according to Terry (or maybe common “misses” by engineers?):

1/ Shear wall horizontal (out-of-plane) offset (i.e., non-stacking wall)

If you are curious and want to open your handy ASCE 7, this is Table 12.3-1 “Horizontal Structural Irregularity Type 4.”

When this occurs, all supporting members need to be designed for overstrength.

(Or convince the client that “stacking the walls” is the way to go.)

(Source: ASCE41-17)

2/ Shear wall in-plane offset (continuous and discontinuous)

Similar to the last one but offset in-plane (“Vertical Structural Irregularity Type 4” per Table 12.3-2).

It is straightforward-ish when the shearwall boundaries come straight down like this:

(Source: Common Challenges in Wood Lateral System Layouts)

It gets tricky when the walls don’t line up nicely (see below) – in this case, the headers and the jambs would need to be designed for overstrength.

(Source: Common Challenges in Wood Lateral System Layouts)

3/ Large diaphragm openings

We should pay attention to how diaphragm forces transfer through drag and chords when we have a large opening.

But what is a “large” opening?

According to the recommendations by FPInnovations (Canadian non-profit wood people, link at the end), if the opening has any of the following, it is considered large:

• Opening depth > 0.15 x (Diaphragm depth)
• Opening length > 0.15 x (Diaphragm length)
• “Distance from diaphragm edge to the nearest opening edge” < 3 x max(Opening depth or opening length)
• “The diaphragm portion between opening and diaphragm edge” exceeds the maximum aspect ratio requirement (i.e., check the diaphragm aspect ratio around each side of the opening against the SDPWS requirements.)

(Source: SDPWS 2015)

If any of the criteria above is true, it is recommended that you should do a detailed analysis (the FPInnovations document has examples on how to do this. Again, link at the end).

4/ Discontinuous chord

Essentially, watch out for corners.

Terry dives into this in more detail in a different WoodWorks document called “The Analysis of Irregular Shaped Diaphragm.” Check out the link at the end to learn more.

(Source: The Analysis of Irregular Shaped Diaphragm)

5/ Cantilever diaphragms

This is allowed by code, but you must meet the SDPWS requirement for “Open Front Structures” (Section 4.2.5.2).

(Source: SDPWS 2015)

6/ Flexible vs. semi-rigid or rigid diaphragm (and global uneven stiffness)

So technically, the code allows you to idealize a diaphragm as flexible if you meet the criteria in ASCE 7 12.3.1.1. This means the analysis is relatively simpler compared to rigid or semi-rigid diaphragms.

However, according to NEHRP Seismic Design Technical Brief No 10 (link at the end), most light-framed wood diaphragms behave semi-rigidly in reality.

The implication is that if you have a building with uneven lateral stiffness on each end, you could have excessive drift on one end even though you technically designed it to meet code.

(Source: ASCE41-17)

So what to do?

You could run a back-of-the-envelope calc treating it like a rigid diaphragm (or cantilevered beam). Based on that, estimate the drift, then stiffen up the lateral system based on educated guess (ahem, I mean engineering judgment).

7/ Combined lateral system

Lastly, according to ASCE 7 12.2.3.3, if you combine different lateral systems (e.g., wood shear wall + moment frame), you should use the lowest R for the direction under consideration.

Don’t need to follow that if you meet all of the exceptions though:

• Risk Category I or II
• Two stories or less
• Light-frame construction or flexible diaphragm

Non-technical Takeaways

Well, that was a lot of technical info. Here are some of my non-technical thoughts:

1. If it feels like you are overthinking things, you are probably not. 90% of the time, we are not designing simple rectangular boxes, and stuff could get pretty complicated, so don’t feel bad. You are not the only one.
2. Being able to identify these challenges early on in the design means you could potentially get the client to change some of them by offering alternatives.
3. And if you are very early (e.g., proposal phase), you might be able to even bake that into your fee, knowing that more effort will be required later down the road.

That is all – thanks for reading!

If you enjoyed reading this, check out the rest of my articles on Back of the Envelope and subscribe maybe?

And be sure to leave a reply to let me know what you think!
(It helps motivate me to keep on writing )

Resources

1. “The Analysis of Irregular Shaped Diaphragm” (link to pdf)
2. “Seismic Design of Wood Light-Frame Structure Diaphragm Systems” (NEHRP Tech Brief) (link)
3. WoodWorks Events Archive with link to pdf (link)
4. WoodWorks Free project support (link)
5. FPInnovations/WoodWorks recommendation for large opening (link)

The post Webinar Recap: “Common Challenges in Wood Lateral System Layout” appeared first on Structural Engineer HQ.

This is a rehash of an article I wrote in my weekly newsletter, “Back of the Envelope” — where I teach you SE-related things in 5 minutes (or less), once a week.

If you enjoyed reading it, consider subscribing at the end of the post to be one of the first to get new emails every Thursday (which I eventually rehash onto SEHQ & LinkedIn a couple of weeks later).

Have you ever set a pencil down vertically on the table to pretend it's a skyscraper, only to then shake that table violently (or blow on it) to knock it down? (you monster)

A standard #2 pencil has a diameter of 0.28 inches, and a length of 7.48 inches — making your skyscraper's aspect ratio equal to about 1:27.

Turns out that they've actually built one of these pencil towers…

(Today's estimated read time = 4 minutes and 30 seconds)

When I first saw that picture (top of this page) of 111 West 57th Street (aka Steinway Tower), I thought it was perhaps Photoshopped.

Except… it's real life, not fantasy.

With a width of 59 ft and a height of 1,428 ft, its ratio is about 1:24 – making it the slenderest tower ever built (so far).

You probably have a lot of questions, such as:

1/ What is the building for?

Super-high-end residential; we are talking about $8M-$66M per unit.

2/ What does the view look like from up there?

(Source: streeteasy.com)

3/ Who designed it?

Arch: SHOP.

Struct: WSP – formerly Parsons Brinckerhoff if you are wondering.

4/ How did they make the structure work?

Thick, high strength concrete shear walls on east & west + core; lots of 70-foot long rock anchors; plus a tuned mass damper at the top.

5/ What's the overturning and uplift?

You nerd.

Just kidding… well, we are structural engineers, so we should run some quick numbers for funsies. In today's issue, we will do a back of the envelope calc on 111-West-57-St for the following:

• Dead load
• Seismic load
• Wind load
• Overturning & uplift

Let's do this.

—

Side note: By the way, obviously they've done tons of serious analyses with this thing (wind tunnel tests, 3d dynamic models etc.); but hey, we are only running calcs with our trusty TI-83 and HP33s, so be easy on me here.

—

Quick-n-dirty dead load

(source: 111w57.com)

Area: The width of the tower is constant at 59 feet, but the length sort of tapers back slightly as it gets taller. I scaled off some plans and got 76′. (So rough floor plate = 4484 sqft).

Slab: I have no idea how thick the slab is, but some videos mentioned 15′ floor-to-floor and 13′ ceiling. So maybe 8″ slab sounds about right (Slab weight = 100 psf).

Wall: 36″ thick at the lower levels and 16″ at the top, so average 26″. Guestimate some wall length (288′) x 15′ tall, divide by area, we get (Wall weight = 313 psf).

Terracotta: Estimate 15 psf but smeared into the floor plate (Cladding = 10 psf).

Miscellaneous: (Misc = 10 psf).

Tune mass damper: 800 ton (TMD = 1600 kip)

(Source: New York Times)

Steel structure: This is the unoccupied structure at the pinnacle. Guessing roughly 10 floors at 30 psf, with an average floor area of 2,000 sqft. (Pinnacle = 900 kip).

All in all, DL = 4484 sqft x 84 floors x (100+313+10+10)/1000 + 1600 + 900 = 165,000 kip.

Put that into perspective in terms of compression on the concrete walls: 165,000 kip / (36 in x 70 ft of wall x 2 x 12) = 2,728 psi

Nice. Seems right.

Quick-n-dirty seismic load

Grabbing some essentials from ATC:

• Sds = 0.304
• Sd1 = 0.096

Make some may-or-may-not-be-true assumptions:

• Risk category = III (not just II because the tower is for the super-rich. Jk? I am guessing it should be higher because it could cause serious damage to its surroundings if it falls down.)
• I = 1.25
• R = 5 (special reinforced concrete shear wall)
• h = 1428’

Not going to bore you with all of the math here but if we follow along with ASCE 7 Chapter 12:

• Cs (Eq 12.8-2) = 0.076
• Cs (Eq 12.8-3) = 0.005
• Cs (Eq 12.8-5) = 0.017

And if I put ASCE 7 words into a formula that you and I intuitively understand (from years of Excel-ing):

Cs = max(min(0.076,0.005),0.017) = 0.017

(I'm definitely not used to seeing this since I am in California… did I make a mistake?)

Seismic base shear is then 165000 x 0.017 = 2,759 kip

(For fun: 2759 / (70’ x 2) = 19.7 k/ft per shear wall.)

Not bad at all. So seismic probably not an issue compared to wind. Let’s see…

Quick-n-dirty wind load

Back to good o’ ATC:

V = 125mph

Make some not-so-accurate-but-probably-ok-enough assumptions:

• Kd = 0.85
• Exposure = C (I have no clue – feels like things are different high up there)
• Kzt = 1.0
• Ke = 1.0
• G = 0.85

Chugging along with ASCE 7 Chapter 26:

• Kz (@100’) = 1.27
• Kz (@500’) = 1.77
• Kz (@1000’) = 2.06
• Kz (@1400’) = 2.21
• qz (@100’) = 43 psf
• qz (@500’) = 60 psf
• qz (@1000’) = 70 psf
• qz (@1400’) = 75 psf

(Phew… almost there)

I am probably butchering this last part following ASCE 7 Chapter 27 but remember, this is just for fun – don't put me in timeout.

• GCp (windward) = 0.8
• GCp (leeward) = -0.5
• GCp (net) = 1.3
• p_net (@100’) = 56 psf
• p_net (@500’) = 79 psf
• p_net (@1000’) = 91 psf
• p_net (@1400’) = 98 psf

And the moment of truth:

Wind base shear = (56×100+79×400+91×500+98×400)x76’ = 9,264 kip

Hey look at that, at least three times higher than seismic.

Alright, I am running out of time (and steam) so going to bump this up a notch to “super-quick-n-dirty.

Super-quick-n-dirty overturning and uplift

OT = 9264 * 1400 / 2 = 6,484,800 kip-ft

T = 6484800 / 58’ = 111,807 kip (uplift from wind)

Net uplift = 111807 – 0.9 x (165000×0.5)  = 37,557 kip

So many digits.

I saw a video that says there are about 200 x 3″ diameter rock anchors. (Here is the video – it's long but pretty good. Watch it at 2x speed for extra fun).

So let's say half of them take the tension.

Tension per anchor = 37557/100 = 376 kip

Stress per anchor = 376 / 7 in^2 = 54 ksi

Not too bad – seems reasonable.

(original image source: Youtube)

And… I am out. (why did I do that to myself?)

I suppose the point is that anyone, including you, could run a back-of-the-envelope calc to see if things make sense-ish.

There are, of course, wind tunnel tests to make the wind load more accurate. Also, in most cases for these towers, drift and occupancy comfort probably govern more than anything else.

But user “MIStructE_IRE” at Eng-Tips said it best:

“…they didn't have ETABS or spreadsheets when they built the empire state! I believe if you can't explain a fancy computer model using a couple of quick calcs and some free body diagrams – then either there's something wrong or the engineer hasn't a clue what they're doing!”

That's all for now. Hope you liked this, and I'll see you next week.

The post Skinniest Tower, Ever appeared first on Structural Engineer HQ.

Today, I am going to talk about “amiability” (aka being friendly).

(Estimated read time: 2 minutes and 31 seconds)

According to Bo Jaquess (current SEAOSD president), “amiability” is number two on the list of “what it takes to excel as an engineer.” (link)

(Which by the way, I talked about #1 here).

Bo says:

“For better or worse, you must work with other people.
The architects, contractors, MEP engineers, welders, and rodbusters are working toward the same goal as you are: a successful project.

And all are under pressure of some kind, just like you are.

Being ‘a pleasure to work with' can be a real factor in your firm landing the next job with that client, as well as your upward mobility within your own firm. Aside from the fact it's just the decent thing to do.”

– Bo Jaquess

Totally agree. Going to riff off on this and add a few of my takes.

To me, being a decent human/engineer also means that:

1/ You are not a jerk

2/ You are not prideful

3/ You don't react to sh*t immediately

Let me elaborate.

— Side note —

By the way, this is a rehash of an article I wrote in my weekly newsletter, “Back of the Envelope” — where I teach you SE-related things in 5 minutes (or less), once a week.

If you enjoyed reading it, consider subscribing at the end of the post to be one of the first to get new emails every Thursday (which I eventually rehash onto LinkedIn a couple of weeks later).

— End side note —

First: You are not a jerk

This is obvious, and most engineers I know are pretty friendly and polite (most of the time).

But the test of character really happens when you are under a lot of stress.

You know, at times when you have deadlines after deadlines, and you haven't slept well in days.

That's when you really got to watch out.

Pay attention to how you react to clients when they call. And pay attention to how you respond to emails.

Do you sense animosity and anger in yourself?

If so, time to step back and pause before you press send (I've done that a few times myself).

You just prevented yourself from becoming a jerk.

2nd: Don't be prideful

Now, this is more common among engineers.

What is prideful?

It means that you think you are better than another person.

“But that contractor just doesn't get it! He butchered my beam and is now asking for help after the fact!”

Yeah I know. I've been there…

But nobody knows everything, so that guy probably knows something that I don't.

If you catch yourself thinking like Scar, maybe it's time to pause and re-evaluate.

Third: Don't react to sh*t

Every day, there will be things that annoy you.

Plan checkers will say that your design is not safe (it is) and does not meet code (it does).

Cost estimators will accuse you of increasing the structural cost between DD and CD.

Architects will ask for your drawings two days before it's actually due.

(These may or may not be based on personal experience )

Whatever it is, sometimes you may feel the need to “react” immediately.

Perhaps fire back with an angry-ish email defending yourself?

Or perhaps in your mind, you pictured yourself as Will Smith (and the person that made you upset is Chris Rock)?

Now before you do anything, just remember this quote from the man on your hundred-dollar bill:

“Whatever is begun in anger, ends in shame.”Benjamin Franklin

(sorry Agent J) 

Again, take a step back, take a walk, take a deep breath, then come back to the situation to see if you feel any differently.

By not cascading the negativity downstream, you might have just made the world a better place (at least a little bit).

And that is part of being amiable.

That is all for now. I know this stuff is not exactly engineering, but it is important to talk about.

Hope you liked it.

If you enjoyed reading this, check out the rest of my articles on Back of the Envelope and subscribe maybe?

And be sure to leave a reply to let me know what you think!

The post Part of what it takes to “excel as an engineer” is to be able to control your emotions appeared first on Structural Engineer HQ.

In today's post, I’ll talk about something I learned recently related to “embedded posts and poles.”

• Quick overview: “nonconstrained” & “constrained”
• What exactly is “rigid floor or pavement”?
• Another option for “chain-link fence”

(Estimated reading time = 3 minutes and 10 seconds)

Overview

You’ve probably seen these two goodies more than a few times throughout your career:

(Source: UpCodes)

They are equations for “embedded posts and poles” from IBC section 1807.3.

We use them to design pole footings for miscellaneous things that cantilever off the ground (e.g., fence posts and flag poles etc.)

The first equation is for “nonconstrained.” Basically, the top of the footing is surrounded by dirt, so it’s free to move horizontally. (By the way, “nonconstrained” is not an actual word in the dictionary, so I am getting the red squiggly line in Word. Why is it not called “unconstrained”… sorry I digress.)

The second equation is for, you guessed it, “constrained.” The code defines it: “lateral constraint is provided at the ground surface, such as by a rigid floor or pavement.”

Using the constrained equation usually allows the footing embedment depth to be shorter than using the nonconstrained equation (potentially saving you or the owner money.)

“Rigid floor or pavement”

The question is then, what exactly is “rigid floor or pavement”?

Some jurisdictions consider it constrained if the top of footing (or post) is surrounded by a concrete slab.

For example, if the post is in the middle of a slab-on-grade, you are good to go.

On the other hand, certain jurisdictions require the top of the footing to be doweled into the slab. Or you can have bars welded to the steel post with enough length to develop the bars into the slab.

Either way, the code is perhaps being intentionally vague so it’s up to interpretation.

Which got me thinking.

These equations have been around for a long, long time. Where did they come from?

Apparently… from tons of research throughout the years (just skim through some of these titles):

(Source: American Society of Agricultural and Biological Engineers)

After some digging, I found this document called ANSI/ASAE EP486.1 Shallow Post Foundation Design, published by the American Society of Agricultural and Biological Engineers (ASABE).

Say what? Agricultural?

Yeah I imagine they probably build tons of posts out there.

My guess is that ASABE studied the researches and compiled their findings into a detailed document. Then the code writers studied the ASABE document and simplified it down to two easy-to-follow equations so we can all enjoy them (if you happen to know the real history – let me know!)

Anyhow. Back to what is considered “constrained.”

Interestingly… check this out.

ANSI/ASAE EP486.1 actually gives you a formula with a nice picture and everything:

(Source: ANSI/ASAE EP486.1)

(In case you are wondering, “table 1” referenced here is the same as IBC Table 1806.2 Presumptive Load-Bearing Values.)

How cool is that?

Essentially, the report considers the constraint resistance to be based on friction + lateral pressure. Sounds reasonable.

If we do a fun run:

• Say you have 5ft x 5ft x 4” concrete slab (Wd = 1250 lbf)
• S = 150 lbf/ft^3
• k_cs = 0.25

The resulting “lateral resistance of constraint” = 400 lbf.

Not bad. And most likely adequate for things you are using the pole foundation for.

Your miles may vary depending on the jurisdiction, but it’s food for thought.

(By the way, for you die-hard enginerds, there is another related document by ASABE that you can dig into: ASABE Meeting Presentation.)

Chain-link fence

Now, suppose you are just using the footings for a chain-link fence. In that case, there is another goodie you can potentially utilize: ASTM 567-14a(2019) Standard Practice for Installation of Chain-Link Fence.

The standard uses a prescriptive method for the footing, which states (just skim through it):

Let me unwrap that for you.

First, the posts have to be spaced at 10’o.c. or less.

Then,

• If posts are 4” or smaller, footing diameter = 4 x post diameter
• If posts are larger than 4″, footing diameter = 3 x post diameter

Now let’s say H = fence height (ft):

• If H is 4’ or less, then footing depth = 24”
• If H is between 4′ and 20′, then footing depth = 24″ + 3″ x (H – 4′)

In other words, if you have the post size, spacing, and fence height – boom! You get footing size.

You can also go the detailed route and run ASCE 7 wind load + pole footing per 1807.3 — but if you are in a hurry and need to get something out quickly to the client, this could be helpful.

(Again, mileage may vary depending on the jurisdiction.)

Alright, and that is all for this week. Hopefully it’s helpful – thanks for reading!

The post Another take on the good ol’ “embedded posts and poles” appeared first on Structural Engineer HQ.

In today's post, I’ll talk about something I learned recently related to “embedded posts and poles.”

• Quick overview: “nonconstrained” & “constrained”
• What exactly is “rigid floor or pavement”?
• Another option for “chain-link fence”

(Estimated reading time = 3 minutes and 10 seconds)

Good morning. Andy here from Back of the Envelope — the place to be for all your p-delta soul-crushing info.

Just kidding. No but seriously. Over the weekend, as I was working on a RISA Floor/3d model for a looming deadline, I repeatedly ran into the dreaded “P-delta diverging” error. It drove me crazy.

So today, I am going to cover:

• What is P-delta
• Why it is important
• What is “P-delta divergence”
• Why it happens and how to resolve it

(Estimated reading time = 4 minutes and 30 seconds)

What is P-delta

Quick refresher.

Say you have a column.

“P” is the axial load applied to the column.

“Delta” is the lateral displacement on one end of the column (caused by whatever reason. E.g., earthquake load or miscellaneous small displacement that just happens in a complex 3d model).

P-delta effect is essentially the extra (aka secondary) moment and shear due to this displacement.

Here is a nice picture from RISA explaining it:

Why is P-delta important?

Because if we don’t account for that secondary moment in the design, we could be under-designing the column.

For example, if your DC ratio for axial load is already high, a little extra moment can put your column into the overstress territory, which is obviously not good.

What is P-delta divergence?

As you can see from the picture above, the secondary shear will cause more lateral displacement, which means… more p-delta effect.

In other words:

Lateral displacement -> secondary shear -> more lateral displacement -> and so on

In practical terms, if we were to do this by hand (don’t):

1/ Apply the load and apply a displacement

2/ Determine the secondary shear caused by p-delta

3/ Apply that shear and determine the new displacement and new secondary shear

4/ Rinse and repeat until the new displacement is so small that it won’t make any more difference — aka the solution converges.

RISA does all this for you in the backend.

But it gets tricky (and soul-crushing) when the solution cannot converge. In other words, the “new displacement” in step 3 is not getting any smaller on each iteration – hence the “P-delta divergence” error.

Why does it happen

Because you screwed up, and your model is broken.

Jk, but actually, yes the model is broken-ish. But how?

If you think about it: What could cause the new displacement to get larger at each iteration?

One reason could be that the column is not restrained by other members. So one end of the column is either moving or rotating out whack — aka “instability.”

Another reason could be because the applied load is causing the member to buckle (remember Euler’s formula from school?). So a solution literally cannot be derived because the column is unstable.

How to resolve it

“How do I get rid of this error asap so I can move on with the rest of my life!” I asked myself.

RISA has a few helpful info to help you debug this (link here and here and here).

They created all of these because, apparently, it is a very common issue.
(“I know I’m not the only one,” sings Sam Smith when he found out that other people were also having p-delta divergence issues).

Long story short, there are 3 steps:

1/ Run the load combination without p-delta and see what happens to the deflected shape (difficulty = easy-ish)

This helped me find my first issue.

The top of my columns did not have rotation stiffness since it’s pinned all around.

The result = whacky deflected shape.

(See the long lines at the bottom? Those are some beams deflecting miles into the center of earth.)

Easy fix — go back to RISA Floor and change the top of columns to “fixed”. This will hopefully resolve the entire issue.

If not, move on to step 2.

2/ Turn on p-delta, but run load combo with only a fraction of the load (difficulty = medium)

In other words, to debug: Create a self-weight dead-load-only combo. Change the load factor to something small like 0.1 to see if the divergence error goes away.

If it does, then it means there is a buckling issue somewhere.

You can potentially find this by looking at the deflected shape in the plan view.

For example, here is a top view of a truss buckling out-of-plane:

3/ Look at the coordinate of the error and hypothesize (difficulty = sucks)

Sometimes the lateral displacement is so tiny that the deflected shape from steps 1 or 2 tells you literally nothing.

And because you know nothing at this point, Jon Snow, the only thing you could really do is locate the coordinate (RISA tells you where it is) and just “look around it” to see if you can figure out what’s wrong.

Luckily for me, after fidgeting for hours with this, I was able to see that one of my columns in the area was not restrained in one of the directions.

Adding a beam where my arrow is showing resolved the issue. And the soul has been uncrushed, for now.

And that’s all – thanks for reading! 

Hope this is helpful (at least maybe someday).

If p-delta divergence ever occurs to you… good luck, but never give up, never surrender. You got this.

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Today, I will talk about a webcast that happened in March called “For the Betterment of the Structural Engineering Profession.”

I’ll give you a brief overview of what went down and tell you my take.

Let’s dive in.

(Estimated read time: 3 minutes and 52 seconds)

The “Vision”

So the good people at CASE, NCESEA, and SEI did an hour-long virtual “town hall meeting” to update us on their vision of the profession and what they are doing to get there.

(Don’t know the acronyms? Don’t feel bad; me neither. ‘S’ stands for structural.)

The official “vision” can be viewed here if you are interested (second page). There are quite a few big words in there (which I am sure they’ve put a lot of thought into), but that also means it takes quite a bit of effort to digest.

To save you some time, this is my shortened version of the vision:

(Really, I think that's what they are trying to say.)

“Key Initiatives”

Now, with that vision, they came up with ten key initiatives. It’s mostly stuff you’ve heard people talking about, like developing leaders, advancing the profession, encouraging resilience, improving mentoring, etc. (you can read the full list here).

They showed the result of a poll they did recently to see which one people cared about the most (I took a screenshot):

The webcast then went over each one, and the panelist talked about what they are doing for each initiative.

It would be nice if they put that down on a website somewhere, but from what I can remember, a lot of it is “we’ve set up committees who have many things down in the pipeline…etc.” (Sorry I am probably not doing them justice here.)

The last part of the session was a round of Q&As which was interesting.

Someone on Reddit (“ma_clare”) copied down the questions being asked (you can read them here) — my impression was that the answers to these questions were pretty general (e.g., “yes we are doing several things related to that” or “the best way for you to help is to get involved”) which are really not that helpful.

All in all, it was an interesting meeting but definitely left a lot to be desired.

My Take

Alright now let me tell you what I think.

First of all, props to these guys for putting it together; you can tell that they’ve spent a lot of time genuinely trying to improve the profession.

However, I see a few potential issues with the way this is approached:

1. It might be a bit too corporate-ish.

Not sure how to describe this but it feels a lot like a politician telling people, “yes we have committees and initiatives to help us reach our goal.” Basically, a lot of words but not really saying a whole lot. I think some specific milestones or roadmaps could help.

2. Too ambitious

The ex-COO/CFO of Instacart, Ravi Gupta, once said this:

My advice to people when they are thinking about instituting a new process is to go to a whiteboard and write down the answer to this question: “If you could only get one thing done this year, what would it be?”. 

If that answer is “institute some new process”, go for it. But if it’s something like “increase market share from 30% to 60%” or “launch this new product that will 2x our TAM”, don’t waste your time on anything else. Just take your best person (up to and including the CEO), make them responsible for solving that problem, and give them everything and everyone they need to make it happen.” 

In other words, everyone should concentrate on one singular focus rather than diverting resources to 10 different things.

Kind of like drawing vectors – 10 vectors going in the same direction get you a lot farther than having arrows going in all directions.

3. They either missed, or purposely avoided talking about “money”?

It is true that many are in this profession because they love the challenge and the work. But if the money can’t keep up, we will have problems keeping and attracting talents down the road (or even now).

In fact, more money also means more resources for the “key initiatives” and which can get us closer to the “vision.”

But how? I have several ideas — we'll save that for another day.

And that’s all for now – thanks for reading!

Let me know what you think in the comments below.

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